Seismic Performance of Exterior RC Beam–Column Joints Retrofitted using Various Retrofit Solutions
International Journal of Concrete Structures and Materials
Seismic Performance of Exterior RC Beam-Column Joints Retrofitted using Various Retrofit Solutions
Gia Toai Truong 0
Ngoc Hieu Dinh 0
Jong-Chan Kim 0
Kyoung-Kyu Choi 0
0 School of Architecture, Soongsil University , 369 Sangdoro, Dongjak-gu, Seoul 06978 , South Korea
Beam-column joints in existing concrete buildings might not satisfy the design requirements for seismic reinforcement details specified in current seismic design codes. Thus, in this study, various retrofit solutions for existing exterior beam-column joints with non-seismic details were developed: head re-bars anchoring, carbon fiber reinforced polymer (CFRP) wrapping, haunch retrofit element, and steel jacketing with various shapes and sizes. To investigate the seismic performance of exterior joints strengthened with the proposed retrofit solutions, seven half-scale exterior reinforced concrete beam-column joints including one control specimen and six retrofitted specimens were fabricated and tested under cyclic loading simulating earthquake loading. The test results showed that the proposed retrofit solutions could partially enhance the seismic capacity of the beam-column joints: steel jackets could increase deformation and load-carrying capacities; steel haunch elements could increase the load-carrying capacity, stiffness, and dissipated energy; and head re-bar anchoring and CFRP wrapping did not significantly effect on the seismic capacity of the beam-column joints.
seismic retrofit; exterior concrete beam-column joints; head re-bars; carbon fiber reinforced polymers; haunches; steel jackets
The moment resisting frame is one of the most widely
used structure systems in reinforced concrete (RC) buildings
subjected to gravity and/or seismic loading. From a
structural point of view, while beam–column joints that connect
main structural members (i.e., beams and columns) are the
weakest components, they are the most complicated
components as they transfer internal forces between the structural
members. In many developing countries (e.g. Korea),
existing concrete buildings constructed before the 1980s
were designed to resist mainly gravity loads. Thus, in the
past, many reinforced concrete buildings in developing
countries collapsed during severe earthquakes (including
those in Costa Rica in 1991
, in Nicaragua
(Kanamori and Kikuchi 1993)
, and in Egypt in 1995
(Suarez et al. 1995)
). It was found that in many cases the
collapse of such reinforced concrete buildings was caused by
inadequate reinforcement details of beam–column joints.
According to a previous investigation
(the Korea National
Emergency Management Agency 2011)
in Korea, concrete
beam–column joints designed without consideration of
earthquake load have certain material and geometrical
characteristics: low concrete strength; plain steel re-bars;
inadequate or no transverse reinforcement in beam–column
joints; and insufficient anchorage detailing leading to lack of
(Santersiero and Masi 2015; Pampanin
et al. 2002; Engindeniz et al. 2005)
. When such concrete
buildings are subjected to earthquake loading, the beam–
column joints (in particular, the corner and exterior joints)
could fail in brittle manner, which might trigger partial or
entire collapse of the buildings
(Rashidian et al. 2016;
Petrone et al. 2016)
. Thus, the rehabilitation of such beam–
column joints is a prerequisite to resist seismic loads.
To develop retrofit techniques for external beam–column
joints with non-seismic details, a number of studies have
Shafaei et al. (2014)
nonseismically detailed RC beam–column joints with steel
angles, which were mounted using prestressed cross-ties.
With the use of the retrofit technique, slippage was prevented
by increasing the joint area of the beam bottom
reinforcement, the plastic hinge was relocated far from the column
face, and the shear strength, stiffness, energy dissipated, and
ductility capacity were also significantly increased up to 50,
120, 220, and 220%, respectively.
El-Amoury and Ghobarah (2002)
investigated the seismic
performance of the joints retrofitted with glass
fiber-reinforced polymers (GFRP). The proposed rehabilitation
schemes consist of two systems: the first system is used for
upgrading the shear strength of the joint with two U-shaped
GFRP layers, and the second system is used for upgrading
the bond-slip of the steel bars. The use of GFRP jacketing
significantly enhanced the ductility and the load-carrying
capacity of the rehabilitated joints (52% higher than that of
non-retrofitted joints). The brittle joint shear failure of the
retrofitted specimens was eliminated, the bonding between
concrete and beam top reinforcement was improved, the
stiffness degradation of the joints was reduced, and the
dissipated energy increased by up to six times compared to
that of non-retrofitted joints.
Del Vecchio et al. (2014) also investigated the seismic
performance of exterior RC beam–column joints retrofitted
with fiber-reinforced polymers (FRP) through an
experimental program carried out on six full-scale test specimens.
From the test results, it was found that the observed
maximum strain of FRP (approximately 1.0%) was larger than
0.4% as design maximum strain and the amount of FRP joint
reinforcement significantly influenced on joint panel
deformations. Based on such test results, analytical models were
also developed to predict the shear strength of beam–column
joints having light reinforcement details retrofitted with FRP
(Del Vecchio et al. 2015; Pantazopoulou et al.
. In addition, Ronagh and Baji (2014) and
et al. (2016
) simulated the shear behaviour of RC beam–
column joints using finite element method (FEM). In the
Del Vecchio et al. (2016
), the anchorage effect of
beam bars and the shear contribution of the FRP
strengthening system were delicately accounted for, and the analysis
results showed a good agreement with the test results in
terms of strength, global deformation, crack pattern, strength
and stiffness degradation, and pinching.
Pampanin et al. (2006)
experimentally investigated the
effectiveness of various haunch retrofit solutions on the
seismic response of the reinforced concrete beam–column
joints. The haunch elements were used including diagonal
axial elements and they were hinged or welded to the steel
plates at both ends to form haunch elements. These haunch
elements were connected to the beams and columns by using
two partially prestressed external rods along with two
anchors, which were directly fastened to the beams and
columns. The shear strength and ductility capacity of the
retrofitted joints were significantly improved, and the failure
mode was changed to flexural hinging in the beam.
The failure modes of the beam–column joints were
investigated and classified into three types according to the study
Lee et al. (2009)
. In Fig. 1, J-failure refers to the
Beam flexural yielding
connection failure before plastic hinges formed at the ends of
adjacent beams, which is associated with low displacement
and ductility capacity. BJ-failure refers to the connection
failure after the plastic hinges develop at both ends of adjacent
beams, and B-failure refers to beam failure in the plastic hinge
regions while beam–column connections remain elastic.
Compared to J-failure, the BJ- and B-failures are more ductile
modes, since BJ- and B-failures involve beam yielding.
Judging from the research results mentioned above and
(Campione et al. 2015; Ruiz-Pinilla et al.
2014; Bansal et al. 2016; Tsonos 2010; Sharma et al. 2014)
it was obvious that such developed retrofit techniques were
effective in enhancing the seismic performance of exterior
beam–column joints. However, the most significant
disadvantage of the previous retrofit techniques was the lack of
practicability. In some cases, the surfaces of the joints
parallel to the beam direction cannot be covered with retrofit
materials because of the existence of beams in orthogonal
directions. The floor space in the upper story could be
considerably minimized because of the retrofit materials; the
durability can be weakened because of the holes drilled
through the beams and columns, and the use of concrete
jacketing could increase the dimensions and weight of the
structures. Thus, the retrofit techniques for exterior beam–
column joints should be designed and constructed
considering the architectural requirements of the existing buildings,
and thus a minimum retrofit is usually allowed by building
owners. In the studies by
Bakis et al. (2002)
and Del Vecchio
et al. (2014), it was indicated that the seismic performance of
the beam–column joints could be effectively enhanced even
by strengthening one side of the joint panel zones.
In this study, for minimum and efficient retrofit of exterior
beam–column joints, various practical rehabilitation solutions
were proposed using internally embedded head re-bars,
carbon fiber reinforced polymer (CFRP) wrapping, steel haunch
elements, and steel jacketing. Thus, to verify the proposed
retrofit methods, experimental studies were performed. In
total, seven half-scale test specimens were constructed; one
was a control specimen and the other six specimens were
retrofitted with different proposed retrofit methods. All test
specimens were then tested under simulated seismic loading.
The structural performance of retrofitted exterior
beam–column joints was analysed in terms of various factors: load–drift
hysteretic behavior, stiffness, dissipated energy, and damping
ratio. The purpose of this experimental program was aimed at
providing better understanding of seismic performance of the
concrete beam–column joints strengthened with different
retrofit materials. An advancement of practical retrofit
solutions for the beam–column joints could be adopted based on
the findings in the current studies.
2. Experimental Program
2.1 Test Specimens
From the study by
, it was found that in the
existing buildings, which were designed to resist mainly
gravity loads, beam–column joints could not perform well
under seismic loading. This was mainly due to the fact that
poor confinement details led to low ductility of the beam–
In Korea, based on the report investigated by the Korea
National Emergency Management Agency (2011), the
beam–column joints of buildings constructed during the
1980s did not satisfy the requirement of seismic
reinforcement details in current design codes. In such existing
buildings, stirrups or ties with 90 hooks and large spacing
were used, and the anchorage of the top longitudinal
reinforcing bars of beams was bent down inside the connection
regions, while the anchorage of the bottom longitudinal
reinforcing bars was bent down away from the connection
regions, which might decrease the strength and deformation
capacity of the connections
(Priestley 1997; the Korea
National Emergency Management Agency 2011)
In this study, to investigate the seismic performance of
reinforced concrete frames after retrofitting, seven exterior
non-retrofitted concrete beam–column joint specimens with
a scale factor of 1/2 were designed, detailed, and constructed
in accordance with the Korea building code
purpose was to simulate the exterior beam–column joints of
existing RC frame buildings constructed in 1980s in Korea.
One of these specimens was used as a control specimen and
the other six specimens were retrofitted by various retrofit
methods. Figures 2, 3, 4, 5, 6, 7, 8 and Table 1 present the
geometries and details of all test specimens. The control
specimen was denoted as J-0, as shown in Fig. 2. In the
figure, a beam with a cross section of 200 mm 9 300 mm
was connected to a column with a cross section of
300 mm 9 300 mm at the mid-height point. The
squareshaped cross section of the columns was determined after
Antonopoulos and Triantafillou (2003)
Shafaei et al.
. It is noted that the purpose of this study was to
investigate the seismic performance beam–column joints
failed in J failure mode, and thus the joint shear strength of a
control specimen J-0 was determined to be less than the joint
shear demand corresponding to flexural yielding of the
(Lee et al. 2009)
. In addition, to avoid column failure
ahead of joint failure, the longitudinal reinforcement ratio of
the column was determined as 3.4%, and the top and bottom
longitudinal reinforcement ratios of the beam were 3.5 and
0.43%, respectively. The expectation of the failure modes of
the beam–column joints is presented in Appendix 1. It is
noted that J-0 has the same material and geometrical
properties as those of the retrofitted specimens, except for the
The aim of retrofitting the exterior joints was to enhance
the ductile behavior after failure of the joints by providing
half joint shear strength (0.5Vc) as residual strength by shear
reinforcement (see Appendix 2). For practical purposes, high
seismic performance of the joints was not considered an aim
of the study. In addition, it should be noted that in test
specimens, only a portion of the joint panel zones was
retrofitted with retrofit materials to minimize the retrofit of
transverse beams and floor slabs in real structures. Figure 3
shows the details of specimen J-A, which was strengthened
with head re-bars having a diameter of 16 mm, a length of
300 mm, and specified yield strength of 400 MPa. In the
figure, eight head re-bars were equally installed into two
vertical lines with 150 mm spacing in the joint panel zone.
For each vertical line, four head re-bars were installed with a
vertical spacing of 50 mm to each other
(ACI 318-14 2014)
The details of specimen J-CFRP are presented in Fig. 4.
J-CFRP was wrapped with two plies of CFRPs and then
fixed with a vertical line of ten CFRP anchors along the
column. The CFRPs used have a length of 1100 mm and a
width of 100 mm, and only two sides of the column section
were strengthened with CFRPs. The presence of CFRPs was
expected to improve the joint confinement and to avoid the
failure of the CFRP anchors
(Ozbakkaloglu and Saatcioglu
2009; Brena and McGuirk 2013)
. It is noted that in this study
only one specimen strengthened with CFRPs was tested. In
fact, many other strengthening layouts of FRPs, in particular,
at the corners between columns and beams were investigated
and showed good retrofit effect
(Del Vecchio et al. 2014;
Mahmoud et al. 2014)
Figure 5 shows the details of specimen J-H, which was
strengthened with haunch elements made by steel plates
(SS400) having a thickness of 16 mm and specified yield
strength of 400 MPa. In the figure, the plastic hinge regions
of the beam and column below the beam, which were
adjacent to the joint panel zone, were three-sided (or
U-shaped), strengthened with steel gutters, and connected
together by a diagonal steel plate to form a haunch. In
addition, two sides of the steel gutters were installed with
anchor bolts HILTI HSL-3 M12/20 (the diameter of 12 mm
and the length of 120 mm). It is noted that haunches are
expected to change the failure mode from J failure to BJ or B
failure. The evaluation of the strength of the steel haunch
elements is presented in Appendix 3. In this study, the b (and
b0) factor (which was used to determine the redistribution of
the shear between the beam, column, and haunch elements in
the joint) is equal to 2.69, indicating that the bending
moments and the shear forces in the beams and columns
could be reduced by up to 30%. Haunches were not installed
in the upper story of the joint to avoid architectural
inconvenience after the retrofit because the haunches could
occupy the space in the upper story.
(c) Photo of J-CFRP
Figures 6, 7, and 8 show the details of specimens J-SJ1,
J-SJ2, and J-SJ3, respectively, which were strengthened with
steel jackets (SS400) having a thickness of 16 mm and the
specified yield strength of 400 MPa. The difference between
the three specimens was the steel jacketing configuration. In
specimen J-SJ1 (Fig. 6), the two opposite sides of the
column have straight steel plates with a width of 100 mm and a
length of 1100 mm. In specimen J-SJ2 (Fig. 7), two opposite
sides of the column are strengthened with U-shaped steel
plates having a width of 100 mm, a height of 200, and a
length of 1100 mm. Specimen J-SJ3 (Fig. 8) was
strengthened with two steel plates placed along the columns with a
length of 510 mm. In addition, two other steel plates with a
s op (v
e 9 )
p b m 0
s m 3
e io 9 9
n m 0
o e m 2
a i a
T tr i
length of 362 mm and a width of 100 mm were also placed
on the column faces above and below the beam. In all three
specimens, the steel jackets were also installed with anchor
bolts (HILTI HSL-3 M12/20). The area of the steel plates
was designed to avoid the failure of the anchor bolts
. The evaluation of the shear strength
contributed by retrofit materials is presented in Appendix 4.
To investigate the seismic performance of the existing
concrete frame buildings constructed during the 1980s in
, a low compressive strength of concrete
(fc0 18 MPa) and yield strength (fy & 300 MPa) of steel
re-bars were used. Table 1 shows the average compressive
strength of each specimen at the loading day, which was
determined based on the compressive tests according to KS
F 2405 (2010). In the table, the compressive strengths for all
specimens ranged from 18.64 to 22.85 MPa.
A tension test of the steel reinforcing bars was also
performed following the guidance specified in KS B 0802
(2013) and 0814 (2001). The specified yield strength of the
D10, D19, D22, and D25 steel reinforcing bars is 300 MPa.
The average yield strengths obtained, fy, of steel re-bars were
406, 375, 383, and 392 MPa for the D10, D19, D22, and
D25, respectively, exceeding the specified yield strength of
300 MPa. Figure 9a shows the typical stress–strain
relationship of steel re-bars for D19 acquired from the test.
The CFRPs used had the average thickness of 0.293 mm,
the ultimate strain of 0.015, the ultimate tensile strength of
3450 MPa, and the elastic modulus of 230,000 MPa as
values provided by the manufacturers. The head anchor bolts
used in specimen J-A has a diameter of 16 mm and the
specified tensile strength of 400 MPa. Figure 9b presents the
tensile test of the steel jacket plates (SS400) used in this test.
In the figure, the yield strength of SS400 was 430 MPa,
exceeding the specified yield strength of 400 MPa.
Hereafter, in the evaluation of the experimental results in this
study, the material strength obtained from the tests is used
instead of the specified strength of concrete, steel re-bars,
and retrofit materials.
2.3 Test Setup
A strong three-dimensional steel frame was assembled to
carry out testing of the beam–column connections subjected
to cyclic lateral loading. Figure 10 presents a schematic
drawing and photos of the test setup. In the figure, the
column of the connection was placed in a vertical position and
the beam was placed in a horizontal position. The test
specimens were considered to be hinged at both column
ends: the top and the bottom ends of the columns were
connected to a hydraulic actuator and strong floor,
respectively. In addition, lateral cyclic loading was applied to the
top part of the columns, simulating earthquake loading by a
500 kN hydraulic actuator. The end of the beam was pinned
by an axial steel link pin connected at its base. A lateral
Link Unit: mm
support system was connected to a reaction frame to prevent
out-of-plane movement of the test specimens.
Figure 11a presents the cyclic loading history used in this
test according to the guide of ACI 374.2R-13 (2013). In the
figure, two cycles were repeated at each drift ratio level of
0.25, 0.5, 0.75, 1, 2, 3, 4, 5%, and so on until failure. The
determination of the drift ratio is shown in Fig. 11b, which
was mainly based on the ratio of the displacement and the
length of the column. In this study, no axial load was applied
on the top of the columns since such application of axial load
increased the joint stiffness
(Pantelides et al. 2002, 2008;
Quintero-Febres and Wight 2001)
. However, further
investigation is needed on the effect of the column compression
force on the behavior of beam–column connections.
Figure 12 presents the location of the linear variable
displacement transducers (LVDTs) attached to the control
specimen. An LVDT TT1 was placed at the upper end of the
column to measure the total lateral displacement. Shear
deformation of the panel zones and the flexural and shear
deformation of adjacent beams were measured by a set of
LVDTs. It is noted that for retrofitted specimens, only LVDT
TT1 was used to measure the lateral displacement. The other
ito0 0.25% 0.5%0.75% 1% 2%
0 2 4 6 8 10 12 14 16
Number of cycles
(a) Lateral displacement history (b) Drift determination on column
Fig. 11 Loading protocol: a lateral displacement history and
b drift determination on column.
LVDTs were not attached to the retrofitted specimens due to
the existence of the retrofit materials on the surfaces of the
columns and beams around the joint panel zone.
Fourteen ordinary strain gauges were attached to the
longitudinal and transverse reinforcing bars of the beams and
columns around the joint panel zones to measure the steel
strains developed during different loading stages. In
addition, ordinary strain gauges were attached to the surface of
the strengthening materials such as headed anchors, CFRPs,
haunches, and steel jackets to measure the material strains.
Figure 13 shows the arrangement of ordinary strain gauges.
In the figure, the column reinforcing bars were gauged at the
location directly above the connection part (strain gauges
nos. 1 and 2). The beam reinforcing bars were gauged at two
different locations: adjacent to the connection part (strain
gauges no. 3–6) and inside the connections (strain gauges
nos. 7–12) to check the yield penetration of beam
reinforcement. Strain gauges nos. 13 and 14 were attached to
measure the transverse re-bar strains of the columns and
beams. Both the displacement and strain data were collected
using the data acquisition system.
2.4 Retrofit Techniques
In this study, seven reinforced concrete beam–column
connections were cast and six specimens were then
retrofitted, excluding the control specimen. The retrofit process
was carried out according to actual construction procedures.
In the case of specimen J-A (Fig. 3), the holes were
predrilled according to a designed configuration, and the work
was carefully carried out to avoid damage of the internal
concrete and steel re-bars. The head re-bars were then
installed through the holes and completely filled with mortar.
In the case of specimen J-CFRP (Fig. 4), before wrapping
the CFRPs, the surfaces of the two column sides were
carefully cleaned to coat an epoxy adhesive layer along the
column length of 1100 mm to ensure the bonding between
the CFRPs and the column surfaces. After this stage, two
plies of CFRPs were wrapped on the prepared surfaces of the
column. The CFRP anchors were then installed in the holes,
which were pre-drilled and cleaned before coating with
epoxy adhesive. It should be noted each CFRP anchor must
be fully impregnated with epoxy adhesive to ensure the bond
between the CFRP sheets and CFRP anchors
In order to strength specimen J-H (Fig. 5) with steel
haunch elements, steel gutters were prefabricated from
plateshaped steel plates and then assembled on the column and
beam surfaces after coating with an epoxy adhesive layer.
Anchor bolts were then installed through the pre-drilled
holes at a length of 120 mm into the concrete column and
beam. Finally, a steel plate was welded on site with the
bottom surfaces of the steel gutters to form a haunch
according to the designed configuration.
In specimens J-SJ1 (Fig. 6), J-SJ2 (Fig. 7), and J-SJ3
(Fig. 8), the steel jackets were prefabricated as plate-shaped
(Figs. 6 and 8) and U-shaped panels (Fig. 7), and the steel
jackets were then assembled on the column faces after
coating with an epoxy adhesive layer. The anchor bolts were
(e) J-SJ1 (f) J-SJ2 (g) J-SJ3
Ordinary strain gauges (longitudinal steel re-bars in columns and beams)
Ordinary strain gauges (transverse steel re-bars in columns and beams) Ordinary strain gauges at the retrofit location to measure strain of steel jackets, CFRPs, head steel re-bars Unit: mm
Fig. 13 Arrangement of strain gauges attached to steel re-bars and retrofit surfaces: a gauge arrangement, b J-A, c J-CFRP, d
H, e J-SJ1, f J-SJ2, and g S-SJ3.
then installed through the pre-drilled holes at a length of
120 mm into the concrete columns. Finally, in specimen
J-SJ3, the free ends of the steel jackets were welded together.
3. Experimental Results and Discussion
3.1 Damage Observations
Figure 14 shows the damage observation and crack
pattern at the end of testing. In the case of control specimen J-0
(Fig. 14a), flexural cracks initially appeared at a drift level
of 0.5% from the bottom of the beam and quickly
propagated to the neutral axis of the beam at a drift level of 1.0%.
Then, at a drift level of 1.5%, thin flexural cracks appeared
at the beam–column interface and spread along the beam
length. With repeating loading cycles, inclined cracks first
began to occur in the joint panel zone at a drift level of
3.0% for the negative direction of loading (in Fig. 15, the
negative axis of drift ratio corresponds to the case where the
beam is subjected to negative bending moment). Then,
numerous inclined cracks in the joint panel zone with a
large width increased as the loading progressed, indicating
shear failure of the joint panel zone. After the occurrence of
the inclined cracks in the joints, the width of the cracks
continued to increase and the cracks in the beams ceased to
increase. Subsequently, at a drift level of 5%, several thin
vertical and horizontal cracks appeared in the joint panel
zone; in addition, flexural cracks appeared at the widened
beam–column interface, which was attributed to the
slippage of the bottom reinforcing bars of the beam. The test
was intentionally terminated at a drift level of 12%, and no
significant crushing of concrete was observed at the end of
testing in the plastic hinge regions of the beam and column;
in particular, no spalling of concrete cover occurred in the
joint panel zone, regardless of the presence of major
inclined cracks with the large width.
The overall behavior of specimens J-A (Fig. 14b), J-CFRP
(Fig. 14c), J-SJ1 (Fig. 14e), J-SJ2 (Fig. 14f), and J-SJ3
(Fig. 14g), which were strengthened with head re-bars,
CFRPs, and steel jackets, respectively, was similar to that of
control specimen J-0. However, in the cases of J-A, J-SJ1,
J-SJ2, and J-SJ3, the number of cracks in the joint panel
zones was significantly reduced; in addition, the joint
inclined crack width in J-A was narrower than that of control
specimen J-0. This was because the use of head re-bars and
steel jackets were effective in arresting the crack
development. Moreover, in the case of J-SJ3, by using the
proposed retrofit method, the slippage of the beam bottom
reinforcing bars was significantly prevented. At the end of
testing (at a drift level of 12%), no spalling of concrete cover
was observed in the joint panel zones. The failure of these
retrofitted specimens, except for J-SJ3, was characterized by
the beam failure in the negative direction of loading with
wide flexural cracks at the beam–column interfaces.
Figure 14d shows the damage and crack pattern of
specimen J-H, which was strengthened with haunch
elements. First, at a drift level of 1.0%, flexural cracks
appeared in the beam; with the increasing drift level,
numerous flexural cracks newly developed but the crack
width was significantly narrower than that of J-0. From the
drift level of 4%, several thin inclined cracks appeared in
the joint panel zone and in the beam for a negative direction
of loading. With repeating loading cycles, at a drift level of
10%, concrete crushing was observed in the beam outside
the retrofitted part accompanying the buckling of the beam
bottom longitudinal re-bars. This was attributed to the
effects of stress concentration caused by the haunch
element, leading to a brittle local failure in the beam and a
sudden loss of load-carrying capacity
(Sharma et al. 2010;
. In addition, the obtained test results
indicated that the use of haunch elements resulted in the
relocation of the plastic hinge region from the interface between
the column and the beam to the beam.
3.2 Load–Drift Hysteretic Relationships
Figure 15 shows the load–drift hysteretic relationships of
the test specimens. In this study, drift ratio (h) was defined as
D/H, where D is the lateral displacement and H is the column
height (see Fig. 11b). The nominal lateral load-carrying
capacities (Pnp and Pnn) of each specimen based on beam
flexural strength were evaluated by using the actual material
strength obtained from test results, where Pnp and Pnn are the
positive and negative nominal lateral load-carrying capacities,
respectively. In Fig. 15, Pnp is less than Pnn because the beams
were designed to resist mainly gravity loads. Pvc was also
evaluated corresponding to joint shear strength.
In Fig. 15, in the negative direction of loading, the
maximum lateral load was greater than that in the positive
direction of loading because the top longitudinal
reinforcement ratio of beams was greater than the bottom longitudinal
reinforcement ratio. In the case of the control specimen J-0,
the maximum lateral load was less than the values of Pnn
predicted based on beam flexural strength (Fig. 15a). This
failure of the joint could be classified as shear failure.
In Fig. 15a, the control specimen J-0 exhibited high
deformation capacity exceeding 5% in load–drift hysteretic
response, even though pinching was severe and significant
strength degradation was observed after the peak load. It is
noted that the pinching effect of the load–drift hysteretic
response leads to low energy dissipation of the specimens
(Liang et al. 2016)
. In the negative loading direction, the
yielding of longitudinal re-bars was not observed;
meanwhile, in the positive loading direction, the longitudinal
rebars had yielded at a drift ratio of 0.92%. After yielding of
the beam bottom longitudinal re-bars, the joint was damaged
with numerous wide inclined cracks and several horizontal
or vertical cracks.
The load–drift hysteresis loops of the retrofitted specimens
were almost the same as that of the control specimen J-0.
However, most retrofitted specimens except for J-A and
J-SJ2 showed maximum lateral loads greater than the values
predicted by beam flexural strength. This indicates that the
joint failed after beam yielding. Thus, in the retrofitted
specimens, the structural behavior would be relatively more
ductile than that of J-0. In the J-SJ series, J-SJ1 and J-SJ3
were more effective than J-SJ2.
3.3 Load–Drift Envelop Curves
Figure 16 compares the load–drift envelop curves for all
RC beam–column joints, which were strengthened with head
re-bars, CFRPs, steel haunch elements, and steel jackets. In
the envelop curves, the peak load, effective stiffness (Ke),
yield drift ratio (hy), ultimate drift ratio (hu), and drift
ductility factor (l) were determined. Figure 17 shows the
definition of the deformation capacity and effective stiffness of a
typical RC beam–column joint. In the figure, Ke is the
effective stiffness which is evaluated as an initial segment of
the backbone curve; he [= 0.75Pmax/(KeL)] is the drift ratio
corresponding to 0.75Pmax in the ascending part of the
envelop curve, where L is the column length, hy is the yield
drift ratio defined at Pmax, hu is the ultimate drift ratio at
specimen failure (where a 20% drop in the descending part
of the envelop curve from the peak load is observed
(0.8Pmax)), and l (= hu/hy) is the drift ductility factor at
specimen failure and is determined by the ratio between the
ultimate drift ratio at specimen failure and the yield drift
ratio. Table 2 shows the effective stiffness (Ke) and the
deformation capacity (hy, hu, and l) in the negative loading
direction of all specimens.
In Fig. 16a, in the negative loading direction, the control
specimen J-0 reached the peak load of -51.75 kN at a drift
ratio of -2.75% and the ultimate drift ratio of J-0 was
5.51%. Meanwhile, in the case of specimen J-A, which was
strengthened with the head re-bars, the peak load and
deformation capacity of J-A were almost the same as those
of J-0. In the case of specimen J-CFRP (Fig. 16b), with the
wrapping of the CFRPs, the peak load and deformation
capacity of specimen J-CFRP were slightly greater than
those of J-0; however, the difference was not considerable.
The peak load of J-CFRP was -56.25 kN at a drift ratio of
-2.78% and the ultimate drift ratio was 5.77%. From the test
results, CFRP wrapping is not effective in increasing either
strength or deformability. However, it is noted again that
only one specimen strengthened with CFRPs was tested and
the strengthening layout of CFRPs used in this test was also
newly developed. Meanwhile, in the studies by
et al. (2014)
Singh et al. (2014)
, the beam–column joint
specimens strengthened with various layouts of CFRPs were
tested and the test results showed that the strength and
ductility of beam–column joints could be increased up to
approximately 72 and 98%. Further research is necessary for
better understanding this behaviour.
Specimen J-H, which was retrofitted with steel haunch
elements, showed greater peak load than that of J-0
(Fig. 16c). In the negative direction of loading, J-H reached
the peak load of -79.31 kN at a drift ratio of -4.66%, but
the deformation capacity was not enhanced. It is clear that
hunch elements are efficient in increasing the strength.
The peak load and deformation capacity of specimens
J-SJ1, J-SJ2, and J-SJ3, which were retrofitted with steel
jackets, were slightly greater than those of J-0. In the
negative loading direction, J-SJ1, J-SJ2, and J-SJ3 reached the
peak loads of -57.27, -54.71, and -56.31 kN at the drift
ratios of -2.81, -2.75, and -2.81%, respectively; the
-2 0 2
Drift ratio (%)
-2 0 2
Drift ratio (%)
(b) CFRP wrapping
J-CFRP 6 8
Ke = 0.75Pmax /θ eL
μ =θ u /θ y
Actual envelop curve
θu Drift ratio (%)
Fig. 17 Definition of drift ratios in idealized backbone curve.
ultimate drift ratios were 6.51, 6.36, and 6.26%, respectively.
From the test results, the J-SJ series using steel jackets are
effective to increase both strength and deformability.
In addition, as shown in Table 2, the effective stiffness and
yield drift ratio of J-H were greater than those of J-0 by
approximately 20.47 and 27.45%, respectively; meanwhile,
the effective stiffness and yield drift ratio of the other
retrofitted specimens were almost the same as those of J-0.
Moreover, the drift ductility (l) of the retrofitted specimens
did not considerably differ from that of J-0.
Fig. 16 Effects of various retrofit applications on load–drift relationships: a head re-bars, b CFRP wrapping, c haunches, and
d steel jacketing.
3.4 Strain Profiles
Figure 18 shows the strain profiles for longitudinal re-bars
of the beams obtained from the test results. In Fig. 18a, the
strains of the top longitudinal re-bars in the beams were the
average values obtained from the steel strain gauges, namely
3 and 4, which were attached to the D22 re-bars and were
located at the beam–column joint interface (see Fig. 13a). In
the negative direction, as shown in Fig. 18a, the strains of
the test specimens increased with the increasing drift ratio
and did not show significant difference. The maximum strain
of J-0 and J-A was -0.00147 at a drift ratio of -2.74%;
thus, the longitudinal re-bar strains of J-0 and J-A did not
reach their yield strain of -0.0015; this means that the beam
top longitudinal reinforcement of J-0 and J-A remained in
the elastic range. Meanwhile, the maximum strains of the
retrofitted specimens ranged from -0.00151 to -0.00170,
exceeding their yield strain of -0.0015. In the positive
direction, yielding of the beam top longitudinal re-bars was
Figure 18b shows the average strains of the beam bottom
longitudinal re-bars, which were obtained from the steel
strain gauges (namely 5 and 6) attached to the D10 steel
rebars and located at the beam–column joint interface (see
Fig. 13a). In the figure, in the positive direction, the
longitudinal re-bar strains of most test specimens reached their
Fig. 18 Strain profiles for longitudinal steel re-bars in beams.
yield strain of 0.0016. After this stage, the retrofitted
specimens J-A, J-CFRP, J-SJ1, and J-SJ2 exhibited significantly
high tensile strain. In the negative direction, except for J-0,
J-H, and J-SJ3, the beam bottom longitudinal re-bars of the
other retrofitted specimens yielded. In general, this data
confirms the effectiveness of the proposed retrofit solutions
and explains the ductile structural behavior observed in the
3.5 Stiffness Degradation
In this study, the secant stiffness of the test specimens at
each point of envelop curves was defined as the ratio
between the lateral load and correlative lateral displacement.
Figure 19 shows the secant stiffness of the test specimens in
the positive and negative directions. In general, the stiffness
of the test specimens decreased with the increasing drift
ratio; such degradation is attributed to the development of
flexural and shear cracking in the beams and in the joint
panel zones. As shown in the figure, in the negative
direction, the retrofitted specimen J-H exhibited higher stiffness
than that of the control specimen J-0; this was attributed to
the enlargement of the joints and the enhanced concrete
confinement in the joints by using the steel haunch elements.
At early drift ratio of 0.35%, J-A also showed higher
stiffness than that of J-0; however, after that, the stiffness of J-A
suddenly decreased. Meanwhile, in the case of the other
retrofitted specimens, at each loading cycle, the stiffness was
almost the same as that of J-0. In the positive direction,
except for J-H, of which the stiffness was higher than that of
J-0, the other retrofitted specimens showed almost the same
secant stiffness as that of J-0.
3.6 Dissipated Energy and Damping Ratio
In this study, the dissipated energy at each loading cycle
was calculated as the area enclosed by a hysteretic loop at
that loading cycle. The cumulative dissipated energy was
evaluated as the summation of areas calculated for each
loading cycle. Figure 20a shows a comparison of the
cumulative dissipated energy for the test specimens
according to the various retrofit applications. In general,
per cycle (Ed)
Deformation (mm) 4
Drift ratio (%) 6 6 J-0
J-CFRP J-H J-SJ1 J-SJ3
J-CFRP J-H J-SJ1 J-SJ3
Drift ratio (%)
Fig. 20 Effects of various retrofit applications on dissipated
energy and damping ratio.
except for J-H, the dissipated energy of most specimens
showed the same trend and no significant difference was
observed corresponding to the control specimen J-0. In the
case of specimen J-H retrofitted with haunch elements, for
the drift ratios below 2%, J-H dissipated almost the same
energy as that of J-0. Then, the dissipated energy of J-H was
significantly greater than that of J-0 by approximately 45%
at a drift ratio of 6.6%.
The damping ratio (n), which is one of the most important
indices characterizing the dynamic response of the
structures, is defined as Ed/4pEs, where Ed is the dissipated
energy per cycle and Es is the elastic strain energy.
Figure 20b shows the damping ratio versus drift ratio of the test
specimens for a negative loading direction. In general, as
shown in the figure, the damping ratios of the retrofitted
specimens were almost the same as that of the control
specimen J-0. At the elastic state, the damping ratios were
around 0.1–0.12, then decreased to 0.036–0.044 at a drift
ratio of 0.77%; beyond 0.77%, the damping ratios slightly
increased to 0.06–0.08.
In this study, seven half-scale reinforced concrete beam–
column joint specimens were designed to simulate the
existing joints in concrete buildings constructed around
1980s in Korea. Six of the specimens were then retrofitted
and tested under seismic loading only and one specimen was
used as the control. The retrofit solutions applied in this test
were head re-bars anchoring, CFRPs wrapping, haunch
elements, and steel jacketing. Based on the results obtained
in this study, the primary findings are as follows:
In the retrofitted joints, except for J-CFRP, cracking
damage (the number and the width of cracking) in the
joint panel zone was significantly diminished by eye.
The retrofit methods developed in this study including
head re-bar anchoring, CFRP wrapping, haunch, and
steel jacketing could partially enhance the strength and
deformation capacity of the beam–column joints:
In the joint retrofitted with haunch elements (J-H),
the lateral load-carrying capacity was significantly
increased up to approximately 53.3% but the
deformation capacity was not enhanced. In addition, in
specimen J-H, the plastic hinge region was relocated
from the beam–column interface to inside the beam.
Retrofitting with steel jackets, the deformation
capacities and the lateral load-carrying capacities of
specimens J-SJ1, J-SJ2, and J-SJ3 were improved up
to approximately 10.7 and 18.1%, respectively.
However, steel jacketing and using steel haunch
elements would be potential for corrosion and it is
difficult to handle the heavy steel plates
et al. 2017)
The uses of CFRP wrapping and head re-bar
anchoring did not effectively improve the strength
and deformation capacity of the beam–column joints
in this test. In the case of J-CFRP, the lateral
loadcarrying capacity and the deformation capacity were
slightly increased but the difference was not
considerable. In contrast, the lateral load-carrying capacity
and deformation capacity of specimen J-A were the
same as those of J-0. Further research is needed for
better understanding this behaviour.
The stiffness of J-H was greater than that of J-0, while
that of the other retrofitted specimens was almost the
same as that of J-0. However, the stiffness degradation
rates of the retrofitted specimens were similar to that of
The energy dissipation capacity of most specimens
strengthened with head re-bars, CFRPs, and steel jackets
showed almost the same trend as that of J-0, but the
dissipated energy of the retrofitted specimen J-H
continued to increase up to 45%. On the contrary, the
application of retrofit materials did not considerably
affect the damping ratio.
This research was supported by a Grant
from the National Research
Foundation of Korea, Grant funded by the Korean Government.
This article is distributed under the terms of the Creative
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and the source, provide a link to the Creative Commons
license, and indicate if changes were made.
According to the studies by
Lee et al. (2009)
et al. (1996)
, failure of the beam–column joints could be
classified into three types: B failure, BJ failure, and J failure.
Figure 1 presents the types of failure mode of the beam –
column joints. In the figure, B failure (Vc/Vu [ 1.8) indicates
flexural yielding of beams undergoing large inelastic
deformation until ultimate rotational capacity without shear
failure in joints; BJ failure (1 B Vc/Vu B 1.8) indicates joint
failure after initial yielding of beam reinforcement; and J
failure (Vc/Vu \ 1) indicates joint failure by shear force
without yielding of beam reinforcement, where Vc is the
joint shear capacity and Vu is the joint shear demand. In this
study, the ratio Vc/Vu of the control specimen (J-0) was 0.73.
The evaluation processes of the joint shear capacity (Vc) and
the joint shear demand (Vu) of the concrete beam–column
joints are presented in Appendix 2.
Pn=Pnp (or Pnn)
Pn=Pnp (or Pnn)
Fig. 21 Force equilibrium in beam–column joint.
Appendix 1: Determination of Failure Modes
of Beam–Column Joints
Appendix 2: Evaluation of Shear Strength
of Beam–Column Joints
In ACI 352R-02 (2002), the shear capacity of the joint (Vc)
is evaluated as follows in Eq. (1):
Vc ¼ 0:083c fc0bjhc ð1Þ
where fc0 is the concrete compressive strength, and bj and hc
are the effective joint width and the depth of the column in
the direction of joint shear being considered, respectively.
The joint shear demand (Vu) is evaluated based on the beam
yield mechanism to check the failure mechanism of the
The nominal lateral load-carrying capacity of the columns
(Pn) is evaluated based on the assumption of yielding in
beam reinforcement at the connection, as shown in Fig. 21.
where Pn = Pnp or Pnn; Vb and Mb are the shear force and
moment developed by yielding of beam reinforcement,
respectively; and Lb and Hc are the length of beams and
Appendix 3: Design Procedure of Haunch
Elements through Retrofit of Beam–Column
Figure 22 shows the bending moment and shear diagrams
of the beam–column joints after retrofitting with haunch
elements according to the study by
. In the
figure, the bending moment (Mbc) and shear in column (Vc0)
at the interface of the joints were evaluated as follows in
Eqs. (3) and (4), respectively.
Mbc ¼ MbðmaxÞ þ ð1
Vc0 ¼ ð1
2 tan a
Fig. 22 Moment and shear diagrams after retrofitting with
: a moment diagram in
the beam and b shear diagram in the column.
b 4 6Lhb þ 3ahb þ 6bL þ 4ab þ I2cIabHb3c þ 32IIbcHaL2Hbbc3 þ 3IbIhcacH2HLcbb2
b ¼ a 3hb þ 6bhb þ 4b2 þ 2K1d2aEccoIsb2 a þ 6aI2bAbc2 þ 2IbIchacb2 þ 3IIbchac2b2 þ 32IbIhcac2b33
b0 ¼ b
Lb tan a
where Mb(max) is the maximum bending moment in the beam; Vb
is the shear force in the beam; b and b0 are the factors expressing
the effectiveness of the retrofit solution by determining the
redistribution of the shear between the beam, column, and
haunch elements in the joint; a and b are the projected length of
the haunch elements on the beam and column, respectively; hb
is the depth of the beam; Ib and Ic are the effective inertial
moment of the beam and column, respectively; a is the angle
between beam and diagonal haunch element; and Kd is the
stiffness of the haunch elements. The shear demand in the joint
(Vu) was evaluated by considering the beam yield mechanism.
Appendix 4: Design Procedure of Head
Rebars, Steel Jackets, and CFRPs for Retrofit
of Beam–Column Joints
In this study, the retrofit purpose of exterior beam–column
joints was to provide the joints with residual strength of half
joint shear strength (0.5Vc). Thus, shear strength (Vs)
provided by retrofit materials should be greater than 0.5Vc.
Vs 0:5Vc ¼ 0:0415c fc0bjhc ð11Þ
In the cases of specimens retrofitted with head re-bars, Vs
is calculated according to
Vs ¼ Avfytðl sin a þ cos aÞ
where fyt is the tensile strength of head re-bars, Av is the total
area of head re-bars crossing an inclined shear plane, a is the
angle between shear reinforcement and shear plane in the
joint panel zone (in this study, a is assumed as 45 ), and l is
the coefficient of friction (= 1.4 for normalweight concrete
In the cases of specimens retrofitted with steel jackets, Vs
is evaluated according to Eurocode 8-3 (2005). It is noted
that only 50% of the steel yield strength of the jacket is used.
Vs ¼ 0:5hc s fyt;dðcot h þ cot bÞ sin b
where ts is the thickness of the steel jackets, bs is the width of the
steel jackets, s is the space of the steel jackets (bs/s = 1, in the
case of continuous steel plates), h is the strut inclination angle, b
is the angle between the axis of the steel jackets and the axis of
the member (b = 90 , in the case of continuous steel plates),
and fyt,d is the design yield strength of the steel jackets.
In the cases of specimens retrofitted with CFRPs, the
contribution of CFRPs to the shear strength is evaluated
according to ACI 440.2R-02 (2012).
Afvffeðsin a þ cos aÞdf
Afv ¼ bntf wf
ffe ¼ efeEf
efe ¼ 0:004
0:75efu ðfor two-sided wrappingÞ
where Afv is the area of the CFRPs, ffe is the effective tensile
stress of CFRPs at ultimate state, df is the depth of the joint
wrapped by CFRPs, a is the angle of CFRPs to the
horizontal axis, sf is the space between two CFRPs plies, b is the
number of wrapping surfaces of members, n is the number of
CFRPs plies, tf is the thickness of the CFRP sheet, wf is the
reduction factor (wf is 0.85 for two-sided wrapping), efe is the
effective strain level of CFRPs, Ef is the elastic modulus of
CFRPs, and efu is the ultimate strain of CFRPs.
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